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Dubai

Harry George Poulos and Andrew J. Davids

Abstract: This paper describes the foundation design process adopted for two high-rise buildings in Dubai, the

Emirates Twin Towers. The foundation system for each of the towers was a piled raft, founded on deep deposits of cal-

careous soils and rocks. The paper outlines the geotechnical investigations undertaken, the field and laboratory testing

programs, and the design process and describes how potential issues of low skin friction and cyclic degradation of skin

friction due to wind loading were addressed. An advanced numerical computer analysis was used for the design pro-

cess, which was carried out using a limit state approach. This necessitated analysis of a large number of load cases,

and the paper describes how the information was processed to produce design information. A comprehensive program

of pile load testing was undertaken, and class A predictions of both axial and lateral load–deflection behaviour were in

fair agreement with the load test results. Despite this agreement, the overall settlements of the towers observed during

construction were significantly less than predicted. The possible reasons for the discrepancy are discussed.

Key words: case history, footings and foundations, full-scale tests, piles, rafts, settlement.

Résumé : Cet article décrit le processus de conception des fondations adopté pour deux tours à Dubai, les « Emirates

Twin Towers ». Le système de fondation pour chacune des tours consistait en un radier sur pieux reposant sur des dé-

pôts profonds de sols et roches de carbonate. Cet article donne les grandes lignes des études géotechniques réalisées,

les programmes d’essais en laboratoire et sur le terrain, et le processus de conception, et décrit comment les problèmes

potentiels de faible frottement de surface et de dégradation cyclique du frottement de surface due aux charges de vent

peuvent être traités. Une analyse numérique de pointe à l’ordinateur a été utilisée pour le processus de conception qui

a été réalisé au moyen de l’approche d’état limite. Ceci a nécessité l’analyse d’un grand nombre de cas de charge-

ments, et l’article décrit comment les données ont été traitées pour produire les informations pour la conception. Un

programme élaboré d’essais de chargement sur pieux a été entrepris et des prédictions de classe A du comportement en

déflexion sous chargement tant axial que latéral ont montré une concordance assez bonne avec les résultats des essais

de chargement. En dépit de cette concordance, les tassements globaux des tours observés durant la construction étaient

appréciablement inférieurs à ceux prédits. On discute des raisons possibles de cet inconsistance.

Mots clés : histoire de cas, semelles/fondations, essais à l’échelle naturelle, pieux, radiers, tassement.

retail and parking podium areas.

The Emirates project is a twin tower development in Figure 1 shows a photograph of the towers just after the

Dubai, one of the United Arab Emirates (UAE). The towers completion of construction.

are triangular in plan form, with a face dimension of approx- The foundation system for both towers involved the use of

imately 50–54 m. The taller office tower has 52 floors and large-diameter piles, in conjunction with a raft. This paper

rises 355 m above ground level, whereas the shorter hotel describes the geotechnical investigation undertaken for the

tower is 305 m tall. These towers are more than double the project and the process used for the foundation design. It

height of the nearby World Trade Centre, which was for- also presents the results of a major program of pile testing

merly the tallest building in Dubai. The office tower is cur- and compares predicted and observed test pile behaviour.

rently the eighth tallest building in the world, and the hotel Finally, some limited data on settlements during construction

tower is the 17th tallest. The twin towers are on an approxi- of the towers are presented, together with the predicted values.

Received 21 April 2004. Accepted 24 November 2004. Published on the NRC Research Press Web site at http://cgj.nrc.ca on

8 June 2005.

H.G. Poulos.1 Coffey Geosciences Pty Ltd., 8/12 Mars Road Lane Cove West, P.O. Box 125 North Ryde, NSW 1670, Australia

and The University of Sydney, Department of Civil Engineering, NSW 2006, Australia.

A.J. Davids.2 Building Structures, Hyder Consulting Pty Ltd., 116 Miller St., North Sydney, NSW 2065, Australia.

1

Corresponding author (e-mail: harry_poulos@coffey.com.au).

2

Present address: Coffey Geosciences Pty Ltd., 8/12 Mars Road Lane Cove West, P.O. Box 125 North Ryde, NSW 1670, Australia.

Can. Geotech. J. 42: 716–730 (2005) doi: 10.1139/T05-004 © 2005 NRC Canada

Poulos and Davids 717

Ground investigation and site Fig. 1. The Emirates Twin Towers soon after completion of con-

characterization struction.

from earlier investigations via a series of boreholes drilled to

about 15 m depth. These revealed layers of sand or silty

sand overlying very weak to weak sandstone. For the twin

tower development, it was clear that this preliminary infor-

mation was inadequate; hence, a comprehensive investiga-

tion was carried out. This investigation involved the drilling

of 23 boreholes to a maximum depth of about 80 m. The

deepest boreholes were located below the tower footprints;

boreholes below the low-rise areas tended to be considerably

shallower. Standard penetration tests (SPTs) were carried out

at nominal 1 m depths in the upper 6 m of each borehole and

then at 1.5 m intervals, until an SPT value of 60 was

achieved. The SPT values generally ranged between 5 and

20 in the upper 4 m, increasing to 60 at depths of 8–10 m.

Rotary coring was carried out thereafter. Core recoveries

were typically 60%–100%, and rock quality designation val-

ues were also between about 60% and 100%.

Figure 2 shows the borehole information along a section

that passes through the two towers. It was found that the

stratigraphy was relatively uniform across the whole site, so

it was considered adequate to characterize the site with a

single geotechnical model. The ground surface was typically

at a level of +1 to +3 m Dubai Municipality datum (DMD),

and the groundwater level was relatively close to the surface,

typically between 0 and –0.6 m DMD. The investigation re-

vealed seven main strata, which are summarized in Table 1

by material descriptions commonly adopted in Dubai.

geotechnical model Test results

From the viewpoint of the foundation design, some of the

In situ and laboratory testing relevant findings from the in situ and laboratory testing were

Because of the relatively good ground conditions near the as follows:

surface, a piled raft system was deemed appropriate for the (i) The site uniformity borehole seismic testing did not re-

foundation of each tower. The design of such a system re- veal any significant variations in seismic velocity, thus

quires information on both the strength and the stiffness of indicating that it was unlikely that major fracturing or

the ground. Consequently, a comprehensive series of in situ voids would be present in the areas tested.

tests was carried out. In addition to standard SPTs and per- (ii) The cemented materials were generally very weak to

meability tests, pressuremeter tests, vertical seismic shear weak, with uniaxial compressive strength (UCS) values

wave testing, and site uniformity borehole seismic testing ranging between about 0.2 and 4 MPa, with most values

were carried out. lying within the range of 0.5–1.5 MPa.

Conventional laboratory tests, including classification (iii) The average angle of internal friction of the near-surface

tests, chemical tests, unconfined compression tests, point load soils, from direct shear box tests, was about 31°.

index tests, drained direct shear tests, and oedometer consol- (iv) The oedometer data for compressibility were considered

idation tests, were carried out. In addition, a considerable unreliable, because of the compressibility of the appara-

amount of more advanced laboratory testing was undertaken. tus being of a similar order to that of some of the sam-

This included stress path triaxial tests for settlement analysis ples.

of the deeper layers, constant normal stiffness (CNS) direct (v) Cyclic triaxial tests were carried out with one-way cy-

shear tests (Lam and Johnston 1982) for pile skin friction clic loading and a maximum stress of up to 930 kPa.

under both static and cyclic loading, resonant column testing These tests indicated that the unit 4 sand deposit had the

for small-strain shear modulus and damping of the founda- potential to generate significant excess pore pressures

tion materials, and undrained static and cyclic triaxial shear under cyclic loading and to accumulate permanent de-

tests to assess the possible influence of cyclic loading on formations under repeated one-way loading. It could

strength and to investigate the variation of soil stiffness and therefore be susceptible to earthquake-induced settle-

damping with axial strain. ments.

718 Can. Geotech. J. Vol. 42, 2005

Avg. elevation of base

Unit No. Designation Material description of unit (m DMD)

1 Silty sand Uncemented calcareous silty sand, loose to moderately dense –3.3

2 Silty sand Variably and weakly cemented calcareous silty sand –8.1

3 Sandstone Calcareous sandstone, slightly to highly weathered, well cemented –26.8

4 Silty sand Calcareous silty sand, variably cemented, with localized well-cemented bands –33.1

5 Calcisiltite Variably weathered, very weakly to moderately well cemented –53.5

6 Calcisiltite As for unit 5 –68.5

7 Calcisiltite As for unit 5 –79.0

Note: DMD, Dubai Municipality datum.

(vi) The CNS shear tests indicated that cyclic loading had measured values from the CNS tests were within and beyond

the potential to significantly reduce or degrade the skin the range of design values for static skin friction of piles in

friction after initial static failure and that a cyclic stress cemented calcareous soils tentatively suggested by Poulos

of 50% of the initial static resistance could cause failure (1988); that range was between 100 and 500 kPa, depending

during cyclic loading, resulting in a very low postcyclic on the degree of cementation.

residual strength.

Figure 3 summarizes the values of Young’s modulus ob- Geotechnical model

tained from the following tests: (i) seismic data (reduced by The key design parameters for the foundation system were

a factor of 0.2, to account for a strain level appropriate to the the ultimate skin friction of the piles, the ultimate end-

overall behaviour of the pile foundation); (ii) resonant col- bearing resistance of the piles, the ultimate bearing capacity

umn tests (at a strain level of 0.1%); (iii) laboratory stress of the raft, and the Young’s modulus of the soils for both the

path tests, designed to simulate the initial and incremental raft and the pile behaviour under static loading. For the as-

stress states along and below the foundation system; and sessment of dynamic response under wind and seismic load-

(iv) unconfined compression tests (at 50% of ultimate ing conditions, Young’s modulus values for rapid loading

stress). conditions were also required, together with internal damp-

Figure 4 shows the ultimate static shear resistance, de- ing values for the various strata.

rived from the CNS test data, as a function of depth below The geotechnical model for foundation design under static

the surface. With the exception of one sample, all tests loading conditions was based on the relevant available in

showed a maximum shear resistance of at least 500 kPa. The situ and laboratory test data and is shown in Fig. 5. The ulti-

Poulos and Davids 719

mate skin friction values were based largely on the CNS Fig. 4. Ultimate skin friction values from CNS tests.

data, whereas the ultimate end-bearing values for the piles

were assessed on the basis of correlations with the UCS data

(Reese and O’Neill 1988) and also on the basis of previous

experience with similar cemented deposits (Poulos 1988).

The values of Young’s modulus were derived from the data

summarized in Fig. 3. Although inevitable scatter exists

among the different values, there is a reasonably consistent

general pattern of variation of modulus with depth. Consid-

erable emphasis was placed on the laboratory stress path

tests, which should have reflected realistic stress and strain

levels within the various units. The values for the upper two

units were obtained from correlations with the SPT data.

The bearing capacity of the various layers for shallow

foundation loading, pu, was estimated from bearing capacity

theory for the inferred friction angles, the tangents of which

were reduced by a factor of two thirds to allow for the ef-

fects of soil compressibility, as suggested by Poulos and

Chua (1985).

Ultimate limit state

The piled raft foundation was designed via a limit state [1] R*s ≥ S*

approach based on Australian Standard AS 2159–1995,

“Piling—Design and installation” (AS 2159 1995). The de-

sign criteria for the ultimate limit state were as follows: [2] R*g ≥ S*

720 Can. Geotech. J. Vol. 42, 2005

Fig. 5. Geotechnical model adopted for design. Eu, undrained Serviceability limit state

Young’s modulus; E′, drained Young’s modulus, v ′, drained The design criteria for the serviceability limit state were

Poisson’s ratio; fs, ultimate shaft friction; fb, ultimate pile end as follows:

bearing; pu, ultimate lateral pile–soil pressure.

[4] ρmax ≤ ρall

[5] θmax ≤ θall

where ρmax is the maximum computed foundation settle-

ment; ρall is the allowable foundation settlement, taken to be

150 mm; θmax is the maximum computed local angular rota-

tion; and θall is the allowable angular rotation, taken to be

1/350 here.

Load combinations

For each tower, 18 load combinations were analyzed: 1

loading set for the ultimate dead and live loading only; four

groups of 4 loading sets for various combinations of dead,

live, and wind loading for the ultimate limit state; and 1

loading set for the long-term serviceability limit state (dead

plus live loading).

Analyses

Conventional pile capacity analyses were used to assess

the ultimate geotechnical capacity of the piles and raft. For

the piles, this capacity was taken as the sum of the shaft and

base capacities. For the raft, account was taken of the layer-

ing of the geotechnical profile and the large size of the foun-

dation, and a value of 2.0 MPa was adopted for the ultimate

bearing capacity. In these conventional analyses, it was as-

sumed that the portion of the raft providing additional bear-

ing capacity had a diameter of 3.6 m (three pile diameters)

where R*s is design structural strength; R*g is design around each pile.

geotechnical strength; and S* is design action effect (fac- In additional to the conventional analyses, more complete

tored load combination). The above two criteria were ap- analyses of the foundation system were undertaken with the

plied to the entire foundation system, and the structural computer program geotechnical analysis of raft with piles

strength criterion (eq. [1]) was also applied to each individ- (GARP) (Poulos 1994). This program uses a simplified

ual pile. The values for R*s and R*g were obtained from the boundary element analysis to compute the behaviour of a

estimated ultimate structural and geotechnical capacities, rectangular piled raft when subjected to applied vertical

multiplied by appropriate reduction factors. In this case, the loading, moment loading, and free-field vertical soil move-

structural reduction factor was taken as 0.6. For the ments. The raft is represented by an elastic plate, the soil is

geotechnical ultimate limit state, the worst response arising modelled as a layered elastic continuum, and the piles are

from the pile–soil–raft interaction may not occur when the represented by elastic–plastic or hyperbolic springs, which

pile and raft capacities are factored downwards. As a conse- can interact with each other and with the raft. Pile–pile inter-

quence, additional calculations were carried out for actions are incorporated via interaction factors. Beneath the

geotechnical reduction factors of both less than 1 (0.6) and raft, limiting values of contact pressure in compression and

equal to 1. tension can be specified so that some allowance can be made

In addition to the normal design criteria, as expressed by for nonlinear raft behaviour. The output of GARP includes

eqs. [1] and [2], a criterion was imposed for the whole foun- the settlement at all nodes of the raft; the transverse, longitu-

dation to cope with the effects of repetitive loading from dinal, and torsional bending moments at each node in the

wind action, as follows: raft; the contact pressures below the raft; and the vertical

loads on each pile. In addition to GARP, the simplified

boundary element program deformation analysis of pile

[3] ηR*gs ≥ S *c groups (DEFPIG) (Poulos and Davis 1980) was used to ob-

tain the required input values of the pile stiffness and pile–

where R*gs is design geotechnical shaft capacity; S *c is maxi- pile interaction factors for GARP and for computing the

mum amplitude of wind loading; and η is a factor assessed overall lateral response of the foundation system (ignoring

from geotechnical laboratory testing. The value selected for the effect of the raft in this case).

η, on the basis of laboratory data from CNS tests, was 0.5. Both GARP and DEFPIG were used for the ultimate limit

The value for S *c was obtained from computer analyses, state, with undrained soil parameters for the wind loading

which gave the cyclic component of load on each pile for cases and with drained soil parameters for the dead and live

various wind loading cases. loading only cases. The pile and raft capacities were fac-

Poulos and Davids 721

Fig. 6. Computed final settlement contours for the hotel tower. Table 2. Computed maximum settlement and angu-

Contour interval: 5 mm. lar rotation ultimate limit state.

Max. settlement Max. angular

Tower (mm) rotation

Office 185 1/273

Hotel 181 1/256

lar rotation serviceability limit state.

Max. settlement Max. angular

Tower (mm) rotation

Office 134 1/384

Hotel 138 1/378

dicates that the main geotechnical design criterion in eq. [2]

is satisfied: that is, the reduced foundation resistances

clearly exceed the worst ultimate limit state loadings. For

both foundation systems, the average ratio of cyclic compo-

nent of load to design shaft resistance was found to be less

than 0.5, thus satisfying the requirements of the criterion in

eq. [3] for cyclic loading.

Table 3 summarizes the computed maximum settlement

tored, as discussed in the “Ultimate limit state” section and angular rotation under serviceability loading conditions

above. They were also used for the serviceability limit state, as determined by the GARP analyses. Although the com-

but the pile and raft resistances were unfactored in this case. puted values are relatively large, they nevertheless satisfy the

design criteria set out in the “Serviceability limit state” sec-

Foundation design tion, above. Thus, the overall foundation systems were as-

sessed to be satisfactory from the viewpoint of

Pile layout serviceability.

The number, depth, diameter, and locations of the founda- Figure 6 shows the contours of settlement for the hotel

tion piles were altered several times during the design pro- tower after 24 months as computed by the GARP analyses.

cess. The geotechnical and structural designers collaborated Similar settlement contours were developed for the office

closely in an iterative process of computing structural loads tower, which had a somewhat different pile layout. It can be

and foundation response. In the final design, the piles were observed that for both towers, the predicted settlements

primarily 1.2 m in diameter and extended 40 or 45 m below showed a “dishing” pattern, with the settlements near the

the base of the raft. In general, the piles were located di- centre being significantly greater than those near the edge of

rectly below 4.5 m deep walls, which spanned the raft and the foundation.

the first-level floor slab. These walls acted as “webs”, which

forced the raft and the slab to act as the flanges of a deep Raft design

box structure. This deep box structure created a relatively During the design process, the effects of raft thickness

stiff base for the tower superstructure, although the raft itself were studied, but it was found that the performance of the

was only 1.5 m thick. Figure 6 shows the foundation layout foundation was not greatly affected by raft thickness within

for the hotel tower; the piles are generally located beneath the range of feasible thicknesses considered. It was therefore

the load-bearing walls. Also shown in this figure are the decided to use a raft 1.5 m thick for the final design.

contours of predicted final settlement, which will be dis- Initially, GARP was used to obtain estimates of the largest

cussed later. bending moments and shears in the raft for any of the com-

binations of ultimate limit state loadings. Subsequently, it

Ultimate limit state: overall foundation was realized that the moments thus computed were likely to

Table 2 summarizes the maximum computed settlement be greater than the actual moments, because no account was

and angular rotation for each tower under the worst ultimate taken of the effects of the stiffness of the structure itself in

limit state loadings as determined by the GARP analyses. these calculations. Therefore, for the final assessment of raft

For the cases here, the pile and raft capacities have been re- moments and shears, the computed pile stiffnesses and raft

duced by a geotechnical reduction factor of 0.6. The calcu- contact pressures were provided to the structural engineer,

lated values of settlement and angular rotation are not who put them into a program for the complete analysis of

meaningful, but the fact that the analysis does not predict the structure and foundation. Although the settlements were

722 Can. Geotech. J. Vol. 42, 2005

generally similar to those computed by GARP, the resulting quency range of interest (up to about 0.2 Hz), dynamic ef-

values of moment and shear from the structural analysis fects on stiffness were minor, and in general the static stiff-

were significantly smaller than those from the oversimplistic ness values provided an adequate approximation of the

modelling of the raft as a uniform flat plate in the GARP dynamic foundation stiffness.

analysis. The range of frequencies was assessed to be lower than

the natural frequency of the soil profile, which was of the or-

Pile design der of 0.7–0.8 Hz. As a consequence, little or no radiation

To enable assessment of the piles from the standpoint of damping could be relied on from the piles, so all the damp-

structural design, the maximum axial force, lateral force, and ing would be derived from internal damping of the soil.

bending moment in each pile were computed by the follow- From the resonant column laboratory test data, the average

ing process: (i) the maximum axial force was computed value of the internal damping ratio was found to be about

from the GARP analyses for the various loading combina- 0.05. Following the recommendations of Gazetas (1991), the

tions; and (ii) the maximum lateral shear force and bending foundation damping ratio was taken to be 0.05 for vertical

moment were computed with the program DEFPIG, allow- and rocking motions and 0.04 for lateral and torsional mo-

ing for interaction effects among the piles but ignoring any tions.

contribution of the raft to the lateral resistance. The overall

group was analyzed under the action of the various wind Seismic hazard assessment

loadings. A seismic hazard assessment was carried out by a special-

It was found that the largest axial forces were developed ist consultant, and for a 500 year return period, the peak

in the piles near the corners and in two of the core piles. A ground acceleration was assessed to be 0.075g. Assessments

number of the piles reached their full geotechnical design re- were then made of the potential for ground motion amplifi-

sistance, but the foundation as a whole could still support cation and for liquefaction at the site. Because of the lack of

the imposed ultimate design loads and therefore satisfied the detailed information on likely earthquake time histories, the

design criterion in eq. [2]. potential for site amplification was estimated simply on the

Combined with the moments developed by the lateral basis of the site geology, related to the shear wave velocity

loading, the load on some of the piles fell outside the origi- within the upper 30 m of the geotechnical profile (Joyner

nal design envelope for a 1.2 m diameter pile with 4% rein- and Fumal 1984). On this basis, the site was assessed to

forcement, as supplied by the structural engineer. To address have a relatively low potential for amplification.

the problem of overstressing of the piles, a number of op- The presence of uncemented sands near the ground sur-

tions were considered, including increasing the reinforce- face and below the water table suggested that the possibility

ment in the 1.2 m diameter piles, increasing the number of of liquefaction during a strong seismic event. The grading

1.2 m diameter piles in the problem areas, and increasing the curves for these soils indicated that they might fall into the

diameter of the “problem” piles to 1.5 m. The second option range commonly considered to be easily liquefied. The pro-

was adopted, and for the office tower, the total number of cedure described by Seed and de Alba (1986) was used as a

piles was increased from the original 91 to 102, and for the basis for assessing liquefaction resistance with SPT data.

hotel tower, the number of piles was increased from 68 to 92. Because of the greater propensity of the calcareous sand to

generate excess pore pressures under cyclic loading, a con-

Dynamic response servative approach was adopted, in which only a small

The structural design required information on the vertical amount of fines was considered, and the design earthquake

and lateral stiffness of the individual piles in the two tower magnitude was assumed to be 7.5. The overall risk of lique-

blocks for a dynamic response analysis of the entire struc- faction was assessed on the basis of the liquefaction poten-

ture–foundation system. For the pile stiffness calculations, tial index defined by Iwasaki et al. (1984). This index

the program DEFPIG was used, with the following simplify- considers the factor of safety against liquefaction within the

ing assumptions: upper 20 m of the soil profile. On this basis, the risk of liq-

uefaction was judged to be low to very low, depending on

(i) Each pile carried an equal share of the vertical and lat-

the borehole considered. Consequently, there appeared to be

eral loads (although this might not have been an entirely

no need to consider special measures to mitigate possible ef-

realistic assumption, it did enable the stiffness of each

fects of liquefaction within the upper uncemented soil lay-

pile within the group, allowing for interaction effects, to

ers.

be evaluated straightforwardly).

(ii) The loadings from the most severe case of wind loading

were considered. Site settlement study

(iii) The loading was very rapid, so undrained conditions

prevailed in the soil profile. An assessment was made of the settlement over the entire

(iv) The pile heads were fixed against rotation to simulate site at various times after the commencement of construc-

the effect of the restraint provided by the raft. tion. This study was undertaken to facilitate the design of

(v) The dynamic stiffness of the piles in the group environ- structural interfaces between various parts of the project and

ment was equal to the static stiffness. in particular the interface between the towers and the po-

To check the latter assumption, approximate dynamic dium structure.

analyses were also undertaken, using the approach outlined The methodology involved the integration of the settle-

by Gazetas (1991), incorporating dynamic interaction factors ment due to each of the towers (considered as distributed

and dynamic pile stiffnesses. It was found that in the fre- loadings) and due to the low-rise structures (considered as a

© 2005 NRC Canada

Poulos and Davids 723

Fig. 7. Computed settlement contours around the site after Table 4. Summary of pile load tests.

24 months. Contour interval: 10 mm.

Test Diameter Length Max. test

Tower pile No. (m) (m) Test type load (MN)

Hotel P3(H) 0.9 40 Compression 30.00

P1(H) 0.6 25 Static tension 6.50a

P2(H) 0.6 25 Cyclic tension 3.25b

P2(H) 0.6 25 Lateral 0.20

Office P3(O) 0.9 40 Compression 30.00

P1(O) 0.7 25 Static tension 6.50a

P2(O) 0.7 25 Cyclic tension 3.25b

P2(O) 0.7 25 Lateral 0.20

a

Initial estimated value; actual value was different.

b

Maximum load in cyclic test = 0.5 × maximum load in static tension

test.

was assumed. A large Excel spreadsheet was developed to

allow the summation of the effects of all 188 circular loads

and two towers assumed in the model. The settlement at 289

points over the site was computed for 6-month intervals after

the commencement of construction. A typical contour plot

series of equivalent uniform loadings) at defined points for 24 months is shown in Fig. 7.

across the site. For each of these loadings, the relationship

between settlement and distance was obtained. The follow- Pile load test program

ing procedure was developed:

(i) For the towers themselves, the settlements were avail- As part of the foundation design process, a program of

able from the GARP analyses for the serviceability pile load testing was undertaken, the main purpose being to

loadings. assess the validity of the design assumptions and parameters.

(ii) The settlements due to tower loadings that occurred at The test program involved the installation of three test piles

points outside the towers were computed with the pile at or near the location of each of the two towers. Table 4

group settlement (PIGS) computer program (developed summarizes the tests carried out. All piles were drilled under

in-house by the first author). This program uses a sim- bentonite slurry support, with steel casing being provided in

plified approach to compute the settlements both within the upper 3–4 m of each shaft. Because of the very large de-

and outside pile groups subjected to vertical loading. sign loads on the piles, it was not considered feasible to test

The PIGS program uses the equations of Randolph and full-size piles in compression, and as a consequence, the

Wroth (1978) to compute the single-pile stiffness val- maximum pile diameter for the pile load tests was 0.9 m.

ues, and the approximate approach described in Fleming Nevertheless, it will be observed from Table 4 that the two

et al. (1992) is used to compute pile interaction factors. compression tests on the 0.9 m diameter piles involved a

The Mindlin equations are then used to compute ground very high maximum test load of 30 MN.

settlements outside the loaded area.

(iii) The loads acting on the low-rise areas were modelled as Test details

a series of uniformly loaded circular areas. The com- Figure 8 shows the test setup for the 0.9 m diameter test

puter program finite layer elastic analysis (FLEA) piles. For the compression tests, the loading was supplied by

(Small 1984) was used to compute the variation of sur- a series of jacks, and the reaction was provided by 22 an-

face settlement with distance from each loaded area. chors drilled into the underlying unit 5 calcisiltite. Each an-

(iv) The rate of settlement over time for both the tower foun- chor had a diameter of 100 mm and a total length of 40–

dations and the distributed loads was calculated on the 45 m. The anchors were connected to the test pile via two

basis of two- and three-dimensional consolidation the- crowns (a larger one above a smaller unit) located above the

ory, allowing for the gradual increase of load with time jacks and load cells. For the tension tests, the reaction was

(Taylor 1948). For these calculations, a coefficient of supplied by a pair of reaction piles 12 m long, with a cross-

consolidation of 800 m2/year was assumed on the basis beam connecting the heads of the test and reaction piles. In

of field permeability tests and the assessed Young’s the lateral load tests, the test pile was jacked against the ad-

modulus values for the various layers. The rate of settle- jacent 0.9 m diameter compression test pile, the centre-to-

ment of the towers was based on the solution for an centre spacing between the piles being 4.5 m.

equivalent isolated surface circular load, 40 m in diame- For piles P2(O) and P(2)H, the cyclic tension tests were

ter, located above a compressible material 60 m deep, carried out prior to the lateral loading test. In each of the cy-

with an impermeable base layer and a free-draining sur- clic tension tests, four parcels of uniform one-way cyclic

face layer (Davis and Poulos 1972). For the low-rise ar- load were applied.

eas, the solutions for the rate of settlement of a strip Four main types of instrumentation were used in the test

foundation were used, to allow for the continuity of piles: (i) strain gauges (concrete embedment vibrating wire

loading. type), to allow measurement of strains along the pile shafts

© 2005 NRC Canada

724 Can. Geotech. J. Vol. 42, 2005

and hence estimation of the axial load distribution: (ii) rod installation of the test piles. The following programs were

extensometers, to provide additional information on axial used to make the predictions: (i) PIES, for static compres-

load distribution with depth; (iii) inclinometers (a pair, at sion and tension tests (Poulos 1989); (ii) Static and Cyclic

180°, for each pile for the lateral load tests), to enable mea- Axial Response of Piles (SCARP), for cyclic tension tests

surement of rotation with depth and hence assessment of lat- (Poulos 1990); and (iii) ERCAP, for lateral load tests (CPI

eral displacement with depth; and (iv) displacement 1992). All three programs were based on simplified bound-

transducers, to measure vertical and lateral displacements. ary element analyses that represented the soil as a layered

For the two P3 piles, 44 strain gauges were used, 4 at continuum and were capable of incorporating nonlinear

each of 11 levels, and extensometers were installed at 8 lev- pile–soil responses and considering the effects of the reac-

els; for the P1 and P2 piles, there were 32 strain gauges, 4 at tion piles. Young’s modulus for the piles was assumed to be

each of 8 levels, and extensometers at 5 levels. In general, 30 000 MPa. The input geotechnical parameters for the pre-

the strain gauges performed reasonably reliably. For the of- dictions were those used for the design, as shown in Fig. 5.

fice piles, only 3 of the strain gauges, all on P3(O), did not The program SCARP, however, required additional data on

function properly, and for the hotel piles, 13 strain gauges cyclic degradation characteristics for skin friction and end

(out of a total of 108) did not function properly: 1 on P3(H), bearing. Some indication of skin friction degradation was

4 on P2(H), and 8 on P1(H). The strain gauge readings were available from the CNS test data, but some of the parameters

generally consistent with the extensometer readings. relating to displacement accumulation had to be assessed on

the basis of judgement and previous experience with similar

deposits (Poulos 1988). It was therefore expected that the

Class A predictions predictions for the cyclic tension test would be less accurate

To provide some guidance on the expected behaviour of than those for the static tests.

the piles during the test pile program, class A predictions of

the load–deflection response of the test piles were calculated Predicted and measured test pile behaviour

and communicated to the main consultant, prior to the com-

mencement of testing. The geotechnical model was similar Compression tests

to that used for design (see Fig. 5), with some minor modifi- Comparisons between predicted and measured test pile

cations to allow for the specific conditions revealed during behaviour were made after the results of the tests were made

Poulos and Davids 725

Fig. 9. Predicted and measured load–settlement behaviour for Fig. 10. Predicted and measured axial load distribution for pile

pile P3(H). P3(H). DMD, Dubai Municipality datum.

load–settlement curves for test P3(H) and reveals a fair mea-

sure of agreement in the early stages. The predicted settle-

ments, however, exceed the measured values, and the Fig. 11. Predicted and measured load–uplift behaviour for ten-

maximum applied load of 30 MN exceeded the estimated ul- sion test on pile P1(H).

timate load capacity of about 23 MN. The corresponding

comparison for office tower test pile P3(O) also revealed

good agreement in the early stages, but again, the predicted

ultimate load capacity of 23 MN was exceeded. Indeed, it is

clear from Fig. 9 that the actual ultimate load capacity is

likely to be well in excess of the maximum applied load of

30 MN.

The fact that the actual capacity exceeded the predicted

value was significant, because the values of ultimate skin

friction used for the predictions were well in excess of val-

ues commonly used for bored pile design at that time in

Dubai.

Figure 10 shows the measured and predicted distributions

of axial load with depth for two applied load levels. The

agreement at 15 MN load is reasonable, but at 23 MN the

measured loads at depth are less than those predicted, indi-

cating that the actual load transfer to the soil (i.e., the ulti-

mate shaft friction) was greater than predicted.

Figure 11 compares the measured and predicted load–dis-

placement curves for the static tension test on pile P1(H) and

indicates good agreement up to about 2 MN. At higher

loads, the actual displacement exceeded the predicted value, Figure 12 shows the values of ultimate skin friction in-

but the maximum applied load of 5.5 MN exceeded the pre- ferred from the axial load distribution measurements for

dicted ultimate value of about 4.7 MN. For the office tower both the compression and the tension tests. These values are

test pile, a similar measure of agreement was obtained, al- derived for the maximum applied test loads and are likely to

though the maximum load in that case was about 7.5 MN, be less than the actual ultimate values. Also shown are the

because the test pile had a larger diameter (700 mm) than ultimate values adopted for the design process, and these are

the originally planned 600 mm on which the predictions in reasonable agreement with the measured values; indeed,

were based. the values used for design appear to be comfortably conser-

726 Can. Geotech. J. Vol. 42, 2005

Fig. 12. Ultimate skin friction values: design values and values Fig. 13. Measured and predicted load–uplift behaviour for cyclic

derived from load tests. uplift test on pile P2(H).

for pile P2(H).

substantially larger (by about a factor of two) than the de-

sign values commonly used in the UAE prior to the project.

It appears that the CNS tests, which were used as the pri-

mary basis for selecting the design values of skin friction,

hold significant promise as a means of measuring relevant

pile skin friction characteristics in the laboratory.

Figure 13 shows the results of the cyclic tension test for

hotel tower pile P2(H). Four parcels of one-way cyclic load

were applied, and for each parcel there was an accumulation

of displacement with increasing number of cycles; this accu-

mulation was more pronounced at higher load levels. The

predictions from the SCARP analysis are also shown in

Fig. 13. Although the predictions at loads of <1 MN are rea-

sonable, the theory significantly underestimates the accumu-

lation of displacement at higher load levels. A similar (and

limited) level of agreement was obtained for the test on of-

fice tower test pile P2(O). It had been expected that predic-

tions of cyclic response might not be accurate, and this

expectation was borne out by the comparisons. Clearly, the the distribution of load along the pile. The approach is very

extent of displacement accumulation under cyclic loading approximate and clearly has been unable to reproduce the

was underestimated in the SCARP analysis. This analysis, real behaviour of the pile under cyclic tension loading.

based on an empirical approach originally developed by Nevertheless, from a practical viewpoint, the important

Diyaljee and Raymond (1982), assumes that the accumu- feature of the cyclic tension tests was that a load of about

lated settlement of the soil at each element of the pile is re- 50% of the static ultimate load could be applied without the

lated to the number of cycles of loading and the cyclic stress pile failing (i.e., reaching an upward displacement of the or-

level at that element. The computed settlements are imposed der of 1%–2% of diameter). However, the tests indicated

on the pile as external soil movements, and the analysis that the foundation rotations under repeated wind loading

computes the resulting accumulated pile head settlement and could be larger than predicted if the piles were subjected to a

Poulos and Davids 727

Fig. 15. Measured and predicted deflection distributions for pile Fig. 16. Measured and predicted time–settlement behaviour for

P2(H). DMD, Dubai Municipality datum. the hotel tower.

uplift load capacity.

Lateral load tests dicted performance of the test piles gave rise to expectations

Figure 14 shows the predicted and measured load– of similar levels of agreement for the entire tower structure

displacement curves for the hotel tower test pile. Both the foundations. Unfortunately, this was not the case. Measure-

test pile and the reaction pile responses are plotted. The ments were available only for a limited period during the

agreement in both cases is reasonably good, although there construction process, and these are compared with the pre-

is a tendency for the predicted deflections to be smaller than dicted time–settlement relationships in Fig. 16 for two typi-

the measured values as the load level increases. A similar cal points within the hotel tower. The time–settlement

measure of agreement was found for the office tower pile, predictions were based on the predicted distribution of final

although the initial prediction had to be modified to allow settlement (Fig. 6), an assumed rate of construction, and a

for the larger as-constructed diameter of the test pile. It rate of settlement computed from three-dimensional consoli-

should be noted that the predictions took account of the in- dation theory. At the time of the last available measure-

teraction between the test pile and the reaction pile. Had this ments, the tower had reached about 70% of its final height

interaction not been taken into account, the predicted deflec- (i.e., a height of about 215 m). Figure 16 shows that the ac-

tions would have been considerably larger than those mea- tual measured settlements were significantly smaller than

sured. those predicted, being only about 25% of the predicted val-

Figure 15 shows the predicted and measured deflection ues after 10–12 months. A similar level of disagreement was

profiles along the hotel tower test pile at an applied load of found for the office tower.

150 kN. The agreement is generally good, although the mea- Figure 17 shows the contours of measured settlement at a

surements indicate a reversal of direction of deflection at particular time during construction for the hotel tower. Al-

about 3.5 m depth, a characteristic that was not predicted. though the magnitude of the measured settlements is far

This characteristic has been observed from other analyses smaller than predicted, the distribution bears some similarity

(for example, Matlock and Reese 1960) and may also reflect to that predicted. The predicted ratio of final settlement at

the fact that the stiffness of the ground beyond about re- T4 to that at T15 is about 0.7, which is a similar order to

duced level (RL) –4 m was greater than assumed in the anal- that measured. Thus, despite the considerable thickness of

ysis. The sharp kink in the measured deflection profile may the raft and the apparent stiffness of the structure, the foun-

also be due to the change in stiffness caused by the transi- dation experienced a dishing distribution of settlement,

tion from a cased to an uncased pile. which is similar to that measured on some other high-rise

structures on piled raft foundations, particularly the

Measured and predicted building Messeturm in Frankfurt, Germany (Sommer 1993; Franke et

settlements al. 1994).

The generally good agreement between measured and pre- The disappointing lack of agreement between measured

728 Can. Geotech. J. Vol. 42, 2005

Ratio of max. Max. Min.

Modulus below modulus to near- settlement settlement % load

Case 53 m (MPa) pile modulus (mm) (mm) on piles

Original calculations 80 1 138 91 93

Case 2 80 5 122 85 93

Case 3 200 5 74 50 92

Case 4 600 5 40 23 92

Case 5 600 1 58 32 92

Fig. 17. Measured settlement contours for the hotel tower. Con- Fig. 18. Sensitivity of computed interaction factors to analysis

tour interval: 1 mm. assumptions. s, pile spacing; d, pile diameter.

tem” investigation of possible reasons for the poor predic-

tions. At least five reasons were suggested:

(i) Some settlements may have occurred prior to the com-

mencement of measurements.

(ii) The time–load pattern may have differed from that as- reality, the ground between the piles is likely to be stiffer

sumed. than that near the piles, because of the lower levels of strain

(iii) The rate of consolidation may have been much slower within the rock below the pile tips is also likely to increase

than predicted. significantly with depth, because of both the increasing level

(iv) The effects of pile interaction within the piled raft foun- of overburden stress and the decreasing level of strain. The

dation may have been overestimated. DEFPIG program was therefore used to compute the interac-

(v) The stiffness of the ground below RL –53 m may have tion factors for a series of alternative (but credible) assump-

been underestimated. tions regarding the distribution of stiffness both radially and

On the basis of the information available during construc- with depth. The ratio of the modulus of the soil between the

tion, the first two of these suggested reasons did not seem to piles to that near the piles was increased to 5, and the modu-

be likely, and the last two were considered to be the most lus of the material below the pile tips was increased from

likely. Calculations were therefore carried out to assess the the original 80 MPa to 600 MPa (the value assessed for the

sensitivity of the predicted settlements to the assumptions rock at depth). The various cases are summarized in Table 5.

made in deriving interaction factors for the piled raft analy- Figure 18 shows the computed relationships between in-

sis with GARP. When the program DEFPIG was used to de- teraction factor and spacing for a variety of parameter as-

rive the interaction factors originally used, it had been sumptions. It can be seen that the original interaction curve

assumed that the soil or rock between the piles had the same (No. 1) used for the predictions lies considerably above those

stiffness as that around the pile and that the rock below the for what are considered (in retrospect) more realistic as-

pile tips had a constant stiffness for a considerable depth. In sumptions. Since the foundations analyzed contained many

Poulos and Davids 729

piles, the potential for overprediction of settlements was were made about the modulus values for soil between and

considerable, as small inaccuracies in the interaction factors below the piles, a much closer match to the measured settle-

can translate into large errors in the predicted group settle- ments was possible. The importance of taking proper ac-

ment (for example, Poulos 1993). In addition, Al-Douri and count of interaction effects in pile group analyses and of

Poulos (1994) indicated that the interaction between piles in allowing for a more realistic distribution of ground stiffness

calcareous deposits may be much lower than that between at depth was therefore reemphasized.

piles in a laterally and vertically homogeneous soil. Unfortu- The Emirates project involved close interaction between

nately, this experience was not incorporated into the class A the structural and geotechnical designers in designing piled

pile group settlement predictions for the towers. raft foundations for two complex and significant high-rise

Revised settlement calculations, on the basis of these in- structures. Such interaction has some major benefits in

teraction factors, gave the results shown in Table 5. The in- avoiding oversimplification of geotechnical matters by the

teraction factors used clearly have a great influence on the structural engineer and oversimplification of structural mat-

predicted foundation settlements, although they have almost ters by the geotechnical engineer. Such interaction therefore

no effect on load sharing between the raft and the piles. The promotes the development of effective and economical foun-

maximum settlement for case 4 is reduced to 29% of the dation and structural designs.

value originally predicted, and the minimum settlement is

about 25% of the original value. If this case had been used

for the calculation of the settlements during construction, the Acknowledgements

settlement at point T15 would have been about 12 mm after The permission of His Highness General Shaikh Moham-

11 months, which is in much closer agreement with the mea- med bin Rashid al-Maktoum to publish this paper is grate-

sured value of about 10 mm than the original predictions. fully acknowledged. Dr. James Apted provided expert advice

The importance of proper assessment of the geotechnical in relation to pile testing. Patrick Wong, Robert Lumsdaine,

model in computing the effects of group interaction has Leanne Petersen, Paul Gildea, Jeff Forse, and Strath Clarke

again been emphasized by this case history. were involved in various aspects of the field and design

work and the subsequent supervision of pile construction

Conclusions and testing. Dr. Julian Seidel carried out the CNS testing at

Monash University, Australia.

The comprehensive investigation and testing program for

the Emirates project enabled the site to be characterized

more completely than is usually possible with many pro- References

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compression tests, static and cyclic tension tests, and lateral dimensional conditions. Géotechnique, 22(1): 95–114.

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